 Kuntz, Triska, Reeves, a car, yes, what's his name, Estes, Estes, Estes is not here, Okay, we were talking about the ends of the beams holding up the required shear loads due to the applied loads on the beam. It's complex, it's not complicated. You have to be careful. You only need one number, that's your H over the thickness of the web. And depending on how slender that thing is, how much tendency it has to buckle or not to buckle, you are given different variables, different things you have to do. It's a little unusual in that the fee changes, the fee doesn't usually change depending on the slenderness ratio on most of the stuff we've been doing. Quick review is, for all provisions except G21, which is getting ready to talk about, fees are .9. So since it's not a .9, I guess it's a 1. The basic strength is 6 tenths of the tensile yield, which is the yield in shear times the area of the web times, and then we'll have a reduction factor on here that will indicate how susceptible the web is to buckling. And then he tells this for webs with less than this squared E over F sub Y breakpoint, fee is a 1, and C sub V is a 1. Now you already told your fee is a .9 for everything. What in the world? Is that Estes back here? Well that sucks. Oh my goodness. We thought you were bailing. And so it's temporarily a 1 for a short period of time where your slenderness of the web is less than this breakpoint depending on your steel. Then it's .9 for everything else. The C sub V changes depending on the slenderness. It actually stays the same from 2.24 times square root of E over F sub Y up to 1.1 times square root of 5 for most cases times E over F sub Y. You just put your H over T sub W and you put the numbers in there and you see where you fall. Then when it gets a little more slender you have another C sub V, same fee. When it gets a little more slender you have a different C sub V and the same .9. The only difference in case of V, we said that for webs without stiffeners and this, that's almost always the case, that's already getting pretty bad. Case of V is a 5 except for T shapes. If you had a beam made out of a T then case of V is a 1.2. Here's what stiffeners kind of look like. They weld plates in there, they weld plates in there, they weld plates in there. And they make little shear panels what they call out of these little boxes and that strengthens things up quite a bit. In that case, case of V changes depending on the distance between the stiffeners and the H. Remember how H is not equal to D? We've used that earlier. So pretty straightforward. We had already started on a W14x90 back on our page 212 until we got off in the pictures and the codes and everything that go with this so we have some idea of what's going on. We had already calculated, we didn't calculate it, we probably never will. Pulled it out of a table on page 1-25 for a W14x90. Page 221A, there it is right there. W14x90, it's not on that page. W14x90, there's your H over T sub W, 25.9. So we know what we've got. All we don't know now is where we fall in the scheme of things. So our first break point was the 2.24 square root of, I think we worked that out before. 54, and since our 25.9 is less than 54, then our web is going to yield. It's not going to have any buckling problems to it at all. So the fee will be 1 and the C sub V will be 1. It also says, as pointed out in the spec note, this is the case for most wide flanges, that specification note to the user, the comment, 16.1-325. So all you do then is you LRFD 25.9 less than 54, fee is 1, C sub V is a 1, and so you get 185 kips out of it. That 2.24, that's right out of these. Well, the 25 point, yeah, that's correct. Let me make sure which one. You talking about the 22.24? Okay, that's on table B1 and B2. Maybe you're right, maybe it is B4. Right, and up there it ought to say something about bending as opposed to compression. Well, actually this, they just bring for the same shape. Okay, well here's a 14 by 90. Let's see, we just check that number if that's the question. All right, here is a W14 by 90. And according to this table, the H over TW is 25.9. And do you say you got another number from someplace else? Okay, the break points on there. All right, let me dig. I got to go back a little because once we pass it up, but they're in here. Oh, that's moment strength. Yeah, I'll be able to throw it down in front of us here. That's for sheer break points, right? Okay, who's got a book with this number right there in it? The number we're looking for? These are compression elements, so these will be your flanges. Okay, so we're looking for the web. Webs of 3.76. And these are, this is for the web being buckling in the long direction. In other words, here's your shape and your web can do this. Oh, that's terrible. Okay, and the flange can do this and the web can do this. Oh, that really hurts. But what we're talking about is the web doing this because you're compressing it from the bottom and then you have the load on the top. So these numbers are for the web, as you said, and they are for the flanges, but they're in the long direction. This is in the direction. I don't know where that's going to be. Well, I do know where it's going to be because it tells us 16.1-68 maybe. 68 sounds, you know, about the right way down the road from where we're at. See, now we're going to talk about things that are in sheer. And we have these magic break points at these different places. 1.1, 1.37. Is that what we're talking about? 2.24. Very top of the page. There's a 2.24. See, now I've got to go in the right section, of course, for webs with less than this number, you get a full blast and you get a full blast. And then other things, now then you've got to reduce your fee and you've got to reduce your C sub V. That's sure. Well, I guess that's really... This is what we just looked at, isn't it? Doesn't look like it's on the same. No, it's not. It is. This is right out of the AISC manual. So you're talking about buckling in a different direction. This one we're talking about buckling that looks like that. So since ours is just not subject to buckling at all, it's just too fat and it's too short, then we get, you know, the full 185 kips out of the thing. Then, for example, so-and-so, so-and-so, uniform load maximum support equal to the reaction of case. So he wants to know, here was our case. Our case was a 2.08 kip per foot beam, 45 feet long. You'll find that back on page 212. That's no problem that we were working. And he's used it for a couple more examples now. Maximum reaction on the end would be 2.08 times 45, half for you, half for you. And that comes out 2.08 times 45 cut in half, 46 kips, so less than our 185, so that's the final proof that this beam will not fail in shear in the web. If it fails in buckling in the web due to bending, that's somebody else's job, although none of them do, because they're too short, too fat. They may fail by flange buckling along the axis of the flange. Now, the Dickens, what is this? W-14 by 90. Ah, look at this. Look at that. He's got a table with it in there. Wonderful. Now he tells me. There's that 185 we just worked on. Okay. Now, along with the web having a tendency to buckle, sometimes you will have a girder and you will want to bolt a T-shape or an angle onto the girder and then you'll bring the beam in and you'll want to bolt the beam to the girder. And because you like the floor height constant all the way across, rather than have the flange stuck down underneath this thing, they will cut this piece out that's called coping. If you do not cope the beam, then this little block of steel right here won't tear out because it'll mash up into the flange and it's got all kinds of strength. You would have to tear the whole flange out. But if the flange is missing, then this little piece is subject to what? Never mind. You see it. Block shear. That's right. It works just like when you tore a piece out of a channel or when you tore a piece out of a wide flange. The strength of this thing is a function of the gross area in shear and the gross area in tension and the net area in shear and the net area in tension and a couple of practical things to make sure that you don't have a whole lot of strength here and while they're developing this strength, this piece is already broken. There'll be a limit on that. But it's nothing new. It works exactly as this one did here. The only difference is instead of having two planes in shear, you only have one plane in shear and you have one in tension as you did here. And in this case, unlike most of the other cases, you could actually have another row of, another column of bolts. They really call those rows. In that case, instead of having a nice uniform stress distribution under here, the stress distribution gets kind of un-uniform. Then you're... What was that? Some kind of a factor change from 1 to 0.5? Where would you... Where was it? No, the shear lag factor would be where you've got something in pure tension and the load is having to come down and get through a smaller section. I don't remember. But I'll tell you where you're going to find it and where you're going to find it. Well, I didn't mean a page number. But that's fine. But I was thinking, well, you're going to find it in block shear. Because that's where we found it in the first place. And you do say it's on 16.1-? All those. Okay, thank you. And what was that thing called? It was... U, that's right. I remember the B.S. Well... But I didn't remember what it was. T or U or what it was. That was the use of B.S. And it's either 0.5 for this case, for this case, or 3 or 4 or 5, and it's equal to 1.0 If you just have a single column of bolts, then this dress is pretty uniform. This is pictures from last year. Here's pictures from this year. There's the beam. There's the use of B.S. for 0.5. There's the use of B.S. for 1. And it's on 16.1-412. Thank you. Here's another one where they're running the thing across a girder. This one they're using and they're bolting it to a column. Here's where they're not doing anything. Just sitting it on a pad. That's the shear that you and I are working with right there. That's the area of the beam. We're trying to make sure that it doesn't buckle up to where we are so far. And there is the... Look at that. I figured it'd be somewhere. That must be... Here we've got the equation. That's right. I know what you're talking about. This is in the commentary where they talk about this. And these are the specs right here. And there's the use of B.S. He does tell you... No, this is Segui, so I don't know. But I'm pretty sure that he explains that use of B.S. is either a 1 or a 0.5 and what in the specifications. Then he refers to the commentary because he says, some people aren't going to believe this and they won't know why. And so you'll go to the commentary to find out why. Yeah, that's true. Everything else is a 1. But this is such a common case you need to know. For instance, if you look further back in the... in the manual, you'll find a whole section on bolting things like this and how strong they are. Because there is... some moment applied to this group of bolts and there is some shear applied. So there are some bending moments on there and you'll see some with one row, two rows, three rows, four rows. So he wants you to do a block shear problem. He wants to know how strong this little piece right here is. These are three-quarter inch diameter bolts. So you're going to have to add the extra 16th for fit and 16th for damage to the whole size that you think you see there. You're going to have to have a tension area down here one and a quarter times the thickness of a W18 by 35 beams web. It's 50 KSI steel. You're going to have a gross shear area of 9, 10 of 11 inches tall times the thickness of the flange. Kind of like that. Then you're going to have to take off one, two and a half... one, two, three and a half holes to get the net area in shear. And I have to take off a half of a hole to get the net area in tension. Then you're going to plug it in this equation right here. You're going to know the net shear and the net tension you're going to have, as you said, going to be a one. Here is your ultimate in tension. Here is your... What is this? There you go. It's the ultimate in shear. That's good. And then there is another equation where he would like you to... Sometimes if the yield gets up here and this piece yields and pulls so far it's so strong that this thing here gets in tension on the back end. He doesn't want that to happen. So he's going to limit this term to the 6 tenths of tension yield times the gross area as opposed to the net area. And then the last term is the same as before. So I'll leave it for you to run out the numbers on that case. Here's a good beam. It's got this supported back there pretty far. We'll be going to develop most of the shear strength in the web. This one, I'm not sure that looks a little short, just looks ridiculous. The shear stresses are going to have a hard time getting up in there. As far as how much it'll take the things we've got so far, this is not a function of that. But we'll get into designing these plates and at that time we'll start making sure that they are long enough back in here so that somebody doesn't give this a try. Here's a side view of something bolted on. I don't know if that's an angle or a T. Yes, I do, because here's the top view. There's the T. T is bolted to the column. And then the beam is bolted to the T. And block shear. You've done it. Deflection is also a limit. Two reasons, really. It's a serviceability requirement. But number one, it can't be so much deflection that something really goes wrong. By that something gets bent someplace else because this particular beam has got a lot of deflection to it. And also if it feels unsafe, then it doesn't matter if it is safe or not because nobody's going to rent it. There's two reasons that you limit those deflections. The deflections are actually a serviceability requirement. There's a deflection due to dead load and there's a deflection due to live load. The deflection due to dead load, you can either live with it if you don't care or you can have them camber it out. Tell me when you put the concrete on there and put the dead load on there how much deflection are you going to get? How many quarters of an inch? Well, then if you'll tell them when you order the beam, say I want you to camber that beam with a three-quarter inch upward camber. And then when you get it, you make sure you don't put it in upside down and then when you put the concrete on it, it'll come out dead straight. So that's good. So the beam is straight before the people come on there. The next thing is the live load. And the live load is not something that you can get rid of on the front end. So that's usually what you're really limiting with these equations. There's that W18 by 35 we were talking about. There's the thickness of the web and I kept talking about the thickness of the web. There's depth of the beam. It is a serviceability state, not of strength. Deflections are always computed with service or unfactored loads. I worried about what the worst single event would do to the thing if it deflects way too much during the one worst thing that could ever happen in the world and almost never does, but we can live with that deflection as long as it doesn't fall down and you've already taken care of that by making sure none of the pieces fall off. Deflection due to dead load plus live load doesn't apply to steel beam. Dead load deflection is usually compensated by cambering. Therefore only the live load deflection is a concern to you and also only the service load. So you'll calculate, when the person tells you there's 220 pounds per foot on the beam, live load, that's the service load, and you're getting ready to factor it. You don't factor it for these things. You just use it as is. Another effect of deflection is ponding. It is possible that you have a flat roof, which a lot of buildings do, and it rains and some of the gutters get clogged up, then you get the water stacking in it. That causes more deflection of the beam. That means there's more place for the water to pond and they have controls on that inside of the specifications also. Cambering that out of there will also do the job. So no matter what happens, you have a parapet like you do around this building. If the thing is full of water, they've got it figured out. So that if all of the places where water could get out due to rain, it won't pull the roof down. And I've seen a building where that happened. It was a little 7-eleven over on 29th Street. Here's your moment diagrams for our most common cases. Simply supported beam, deflection in the middle, 5WL 4th over 384EI. That W is going to be a service unfactored load. You're going to make sure it's within somebody's idea of what reasonable is. I should have mentioned that back here. There's a reasonable table that different people feel is appropriate. Maximum live load deflection for a roof supporting a plaster ceiling. In other words, if you have plaster up there, it can crack. So they restrict the deflection pretty severely. If it's some other kind of ceiling like these are, I don't know if they ever... I guess it's a plaster thing. Church is probably. But usually they'll have something like this so that you won't crack it if you have more deflection. And if it's not supporting the ceiling at all, then you probably don't care. And they'll give you different things. They say maximum live, they want this much. Maximum dead end live. Somebody thinks that's appropriate. Max snow or wind load deflection is appropriate. These things come out of different places. He's just telling you typical deflection. And if they have a load on your beam that you say, I don't have one of those, then you go back to your 345 work and find the maximum deflection using one of the methods that you were taught. Put a unit load on it, put the real load on it, integral MMDX over EI. Works like a champ. So for example, he's got a service load of 500 dead, live 550. He weighs 35 pounds a foot. If he's going to be picky about it, he may go ahead and add that 35 pounds on there somewhere. Probably this already includes it. This is more convenient to express it in inches and feet, so he's just going to go ahead and change everything the inches before he gets started. He's been told that the live load deflection of L over 360 has been called for. He wants to know if it's okay. There's deflection equations. 5W dead, live 4th over 384 EI. We'll tell him what he's going to have to camber out or live with, and he may not care. Not care. Live load deflection, the only thing he does is he changes the load from 5 to 550.678. But the live load deflection has been restricted in this design to L over 360. 30 feet long, 12 inches a foot. It's a 1-inch as permitted, so his live load deflection is not excessive. If you're not told what that number is, then obviously it's either somebody else's problem or they don't care. Now designing a beam, going the other way. Old Pop Quiz. Good one, I might use that Friday. Yeah, Friday. Here's a different one. One of them is 10 inches deep, one of them is 8 inches deep. This one shows the plastic neutral axis here, why? Area above is equal to area below the plastic neutral axis. Where's the plastic neutral axis on this one? At the top, above here? Below here, how much? Yes, what do you mean yes? 1 inch, look at that, you're absolutely right. That's because 10 minus 1 is 9 and 8 plus 1 is 9 and that's the plastic neutral axis. And Z for that one would be 8 times 1 times 1.5 plus 1 times 1 times 0.5 plus 9 times 1 times 4.5, good. So it would be Z and M sub yield would equal to F sub Y times Z and M sub N would equal to what? Don't everybody speak at once? Very rude. Just you. M sub E B F sub Y. M plastic, that's right, M yield, M plastic. Thank you, I'll use that. I'll go with M sub P. M sub N is here off the hook. Better symbol, M sub nominal, how much? 10? Same, that's a nominal strength. M sub plastic. It was just hard to get that out, wasn't it? And then M sub U has to be less than or equal to what? Resistance factor of how much? 0.9 times M sub nominal. A lot of terms, a lot of terms. Alright, so designing one now. Where did my word design go? There we go. Beam design, you're going to have to try and pick one that's got all of these things that are going to work in your favor. Going to have to have enough plastic moment. Going to have to not laterally, torsionally buckle. Going to have to not buckle in the flanges. Not going to buckle in the web. Commonly, he suggests that you compute the required moment strength so you know what you're being asked for. You do that by factoring your loads and getting the maximum moment in the beam. Select a shape that satisfies this strength requirement. It can be done in two ways. You can guess. Just go to the middle of page 100 through 600 and just pick one. I guess you could. And then if it's too strong, pick a lighter one, or vice versa. Or, I like this, or better, use the beam design charts in part three of the manual. They're on page 399 through 142. You've seen them. They look just like this. They've been doing it forever. And that's all they've done is they've written them down for you on these pages. You only see this little bitty snapshot on any given page. But they're all there. And after you pull that beam, check for the shear strength. It's very seldom going to control. Then you check for the deflection. So he's got a regular old wide flange, fit KSI steel, simply supported. Five and a half kip per foot live load. Probably has some dead load of its own here. Oh, I like this. Why do I like this? No lateral torsional buckling. So I know very good and well that the full plastic moment is going to be able to be used in this shape. And he hasn't said it's compact, but I'll probably either pick a compact or I'll pick one out of a table that includes them being non-compact, if that's the case. So I got what I mentioned before. You're going to write down four things. Imsa plastic, lateral torsional buckling, flange local buckling, web local buckling. You'll make sure everything there fits. Now we add the term delta. We know the web's not going to buckle for anything in the book. If there's flange local buckling, we'll have to be a little careful. It's not going to laterally torsionally buckle, so we're looking for MCP. First, he suggests we get our loads. 1.2 dead plus 1.6 live. He says I'm going to just assume the beam has no weight. It's very unlikely the beam I pick will be perfect. And when I throw its own weight on there, it will be too small. So let's assume it doesn't weigh any. That weighs very little. Probably maybe 30, 40, 50 pounds a foot out of 4,500 pounds per foot. It's not likely that it won't still work. We'll check it. Plus 1.2 dead plus 1.6 live, 7.2 kit per foot. This has a maximum moment WL squared over 8. What was that 5 over 384 thing a while ago? What was that for? Deflection. That's right. This is for a moment. It's equal to 1.8 to 7.2 kit per foot, 30 feet squared, 810 kit feet is required. You've got to come up with a fee subending M sub nominal. This will be M sub plastic. It'll give me that number. Assume it's going to be compact. Compact shape with full support. Nominal is equal to plastic. M sub Y, Z sub X. Same thing we just did on that T beam a minute ago. From fee sub B, M sub N is greater than your request. There's your fee sub B, there's your M nominal. That has to exceed your request. Solve for the required Z of your shape. Z is M sub U over fee sub B. F sub Y, 216 inches cubed. If you say too fast, too fast. I can take a look at it. But there's nothing new. I mean, we've been doing this now for weeks. So I need a beam that's got a 216 cubic inches on it for Z. And I got a really nice thing called a Z table. On page 3-24. I need 216. No, no, no, no, no, no, no, no, no, no, no. Look at this. They're grouped in sets. If you need a 190, this one will work. But the ones above it, you'll find this one is the lightest. That'll give you the 190. It's the lightest one out of this whole set. 120, 86, 83, that's the lightest one. But it doesn't matter because that's still not going to work. 216, 216, here we go right here. A W21 by 93. A RW21 by 84 is lighter. That's what you and I would consider the answer for this problem. The real world, if you say I can't get one, unless I want to wait six months for it, then I'll take what I can get. But that'll save me a chunk of change. Now, let's just say I had chosen W14 by 99. I wouldn't, of course, because it's not the lightest. But let's just say it was the lightest. Then probably what I needed was 172 for Z. To go figure out what this person thinks the plastic moment is, we would multiply Z times 50 KSI times 0.9 divided by 12 inches and a foot. And he ought to give us 649 KIP feet. 6, 6, hmm. Maybe just a little round off error there. Let's check this one here. It's got a, I'm looking for. Oh, that's with the, that's with, that's a 14 by 90. Oh, that's a 14 by 90. Thank you, thank you. There's a 14 by 90. Let's check it. When I work it out, it comes out 589. Geez, man. What's wrong with this? Oh, that does sound bad. What does that subscript mean? Flange problem, flexure problem. That's right. So this thing here is not really going to work for us. We're never going to get to 173 times 50. So you're not going to be able to really use this table. You can use it, you know, but if you happen to run across an F on the one you choose, then the Z isn't giving you the right answer because you can't reach that Z. The flange will buckle first. However, remember we already mentioned, they'll tell you the truth somewhere. Here's the truth for a W14 by 99. Not 649, but 646. For the 90, not 589, but 573. There's your fees to be mpx that you can really have. That takes into account both the Z and the local buckling of the flange. So what we really should have done rather than following Segui down the road like a bunch of lemmings, rather than getting the Z number, we should have just put his moment. His moment was, the moment required was 810. We should have just gone on here for 810, 810, 810, 810, 810. We still got the same beam because it didn't have an F on it. But that one takes care of that. That takes care. It's adjusted for flange local buckling. This is not adjusted for flange local buckling. There's no way to adjust Z. Z is Z. That's it. I don't know if some of you noticed or you were using a 13th edition, just because you couldn't afford a new one. If you are, these two numbers are reversed. I don't think that'll ever matter to you. So we've picked a beam. I like it. People can do that. Any mentions here about non-compact shapes and why you shouldn't do what I just did, you ought to just go ahead and use the moment if you're going to use those Z-tables. Now the beam design charts are also available out there. Basically speaking, they'll already have the fee in there because you need that. There's your L plastic. Your L when the radius of gyration kicks into Timber Schinkel's equation. There's an open dot. That's a solid dot. You can tell where you're at. This is a typical view for a page. You just get a little piece of a given beam. All of these curves are generated with a C-CB of no Christmas present for you. Reason being, he has no idea what your moment diagram looks like and the shape of the moment diagram is what tells you if you get better than Timber Schinkel's equation. So that's a limitation to what you see written on the page. We can get around it, but we'll have to get around it. One thing is when you do use C-CB other than one and somebody tells you you can multiply this number times 1.3 and it goes right there, that's fine. But when it goes right here and he tells you you can multiply it by two, don't ever let it go past in plastic because that's where the beam's going to fail, plasticly. Here are some typical numbers. I don't remember what he was using this for. Maybe just to demonstrate. This is a W24 by 84. I found that number by going in the Z-tables and finding what his Z was. Well, I actually just looked up his piece of BM sub-X, it's 840. And then it drops down to and past that point I just tracked it page after page after page after page until I found an open dot and that M's, a piece of BM sub-R was 515 and then Tim Schinkel's equation took over. For that beam I looked in the Z-table. You can find L sub-P and L sub-R on this page, Z-table, so I listed L, P and L-R. And I'm saying that as an example here was your request, 748 kit feet coming in this way with a given unbraced length of 6 feet, bingo. Anything above that or to the right of that will do the job. So I looked at this one, but it was dashed. So it means it's not the lightest. That's a W18 by 96. Then I went on up from this point and I ran across a W24 by 76 and that was the one that I chose right there. There was another one above that W24 by 84 which, you know, was heavier. So there's not much reason to go to one that's heavier. Any curve to the right or above is stronger. Charges are entered with the unbraced length and the required strength. Anything above the right gives you an acceptable beam. If you run into a dashed curve, there's a lighter shape right around there. Go dig it out. His example 510, the required design strength was 810 and there was continuous lateral support. I'm sure it was saying anything different from what you and I were saying. Read all this stuff and I may have missed something. Now, this is where I showed you. This is limited by Z. This is limited by flanged local buckling and so he's just going to show you this. That's what you'll see in the book. Here's how you can track down somebody. Follow them home so that you can see how strong they are. You'll find out from your Z-tables how much moment he comes in plastic and you just track him down. He comes out here at about 11 feet. So you ought to find him coming in on the next page around 11 feet at around 750 feet a moment and you just keep on tracking it down. So in other words, if you're given on an exam how strong is a W24x103 if it's 40 feet long, finding a W12x103 can be a real challenge. There's a lot of 40 feet down in here and then you start looking and you start looking and then when you go through all the... There it is right there actually. But if you'll track it down and follow it this way, you're guaranteed to be able to use these tables. So he's got a beam shown. Got to support two concentrated live loads of 20 kips each. Here's his moment diagram. Here's his C-sub-B. C-sub-B is M-M-M. They're all equal so C-sub-B is going to be a one. Got to temporarily ignore the beam's weight. Got to give you 192 kip feet of required ultimate moment. And he says from the charts what you need, 192. What's the unbraced length? 24 feet. 24 feet. 192. Zip, zip. W10x54. Oh, it's not the lightest. Oh my goodness. Gee, I'm really sorry about that. Hope your company didn't go into bankruptcy for that. No, it didn't. You just didn't get the job. What is the lightest? W12x53, two pounds per foot, lighter. And I'd say how do you know that, but I know how you know it because you saw the green line and kept on going up there. You next time. Yes, sir? It's going to be on the syllabus, right? That's correct. It's all going to be on Friday? Yeah, you're going to know that's true. I didn't look and see. Have we not gotten into that yet? It won't be on the quiz. Oh, howdy. Can I ask a question? Yeah, as long as I can keep on getting out of here. What you got? If you can't answer it. I was just going over. Yeah? You kind of threw me off. So any questions? There's no block share. It needs to be required. Right. Well, they're just saying somebody else has been assigned that problem. He was told to do that. So you're not responsible for checking that. If it says block shares, you can omit block share for this problem. No, no. Then I think your real question is, why did they say that? Did they say that because they could look at it and say it's not necessary that we check this? Absolutely not. You always got to check it. Now then they were told, you were told in that problem that you did not have to check block share. On some of our early problems, you were told that because you hadn't yet been shown how. It is welded in just two sides and no block share. Well, that's true. If it's welded on both sides and welded on the end, there's nothing, there's no little block that could pull out. You don't care about block share. Block share is a phenomenon where you have bolt holes.